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Design Guide 1 (Second Edition) Revisions and Errata List The following revisions and errata have been superseded by new editions of the original design guide. Download the newest edition of the design guide and the corresponding errata for the latest information. Also, included are two AISC seismic-related standards: the 2010 AISC Seismic Provisions for Structural Steel Buildings(ANSI/AISC 341-10) and the 2010 AISC Prequalified Connections for Special and Intermediate Steel Moment Frames for Seismic Applications (with Supplement No. 1) (ANSI/AISC 358-10 and ANSI/AISC 358s1-11).
I have a question regarding the AISC Seismic Design Manual (Which references the 13th edition ASIC Manual).I am looking at the beam to column connection requirements for an Ordinary Moment Frame. The example 4.4 in the book uses a W18x40 beam connecting to a W12x35 column. During the check for column panel zone shear they use a value for Mu= 77.2 ft-kips. The example states that this is given in example 4.3. I can't seem to figure out how this 77.2 ft-kip value is derived. Especially since the connection is suppose to be capable of developing 1.1RyMp of the beam. Where does the 77.2 ft-kip moment come from?I must be missing something because this doesn't quite make sense to me.RE: AISC Seismic Design Manual.
OK, I was trying to design the panel zone shear for 1.1RyMp as show in section 11.2a. I guess this doesn't apply for the panel zone then.I am trying to help a fabricator figure out whether or not I need doubler plates for some columns. The project was design by another engineer who is requiring the connection be designed for the full capacity of the beam.
Absent the loads from the frame analysis (from above), I guess I am suppose to use the full capacity of the beam or column (which ever is less).This seems a bit harsh considering since I suspect the beam and column sizes were developed for drift vs strength. RE: AISC Seismic Design Manual (Structural) 23 Mar 11 14:36. With the SMF you would have to then comply with the prequalified connections at the back of the manual.which I do not believe we do. For one thing the column sizes do not comply with the local buckling requirements of the SMF.Like I said, I was just trying to help a fabricator figure out whether or not they needed to estimated doubler plates for the connections. From the information given, you do need to figure that doubler plates are required.
Which is going to be a very pricey requirement.For the structures I design, I usually opt out of AISC 341 as allowed by the code.RE: AISC Seismic Design Manual (Structural) 23 Mar 11 14:50. Nutte,I agree that they requirements goes above and beyond those for OMF. There is nothing wrong with that.
Just that it will end up costing a ton of $ to implement if the sections sizes were based upon serviceability and not strength.At this point, I was only asked to help with figuring out if doubler plates were required or not. Based upon the requirements given on the drawings, they are. I have relayed this information to my client. I have also given them a quick run down with regards to designing the connections for the full capacity of the beam.
How they proceed with this information and quote the job is up to them.If they have a good sales team, knowing this information should not hurt them when trying to win the job.
I'll preface this by mentioning I have never done a seismic design.But I recently attended a seismic design conference and, if I remember correctly without looking at my notes, the overstrength factor has nothing to do with lateral irregularities. It has to do with the fact that materials (concrete and steel specifically) have more capacity than specified.
The overstrength factor takes this into account, as you want your 'fuses' to be ductile; i.e., you want them to yield. Consequently, you need to accurately estimate the strength.There's a distinct possibility I pulled that out of my rear and that the factor I'm talking about is called something else. In any case, this is the type of topic a lot of people will probably chime in on. RE: When to and when not to use the overstrength factor.
It is required where it is explicitly stated in the code.ASCE 7 says, 'where specifically required' so you simply have to keep your eyes open for references to section 12.4.3.In ASCE 7-05 look here for a few:12.10.2.112.13.6.412.13.6.512.13.6.6Also, in AISC's 341, there are numerous reference to use of the overstrength factor in special load combinations - they term it 'amplified seismic load'. Here are a few sections in 341:Part I8.4a9.6(a)9.7b.(1)(a)13.2c14.4I'm sure I've missed a few. The key is to iread/ through the codes prior to design and understand all the sections that apply to your particular case. RE: When to and when not to use the overstrength factor. (Structural) 10 Jul 08 19:53. Frv is correct.I was at the steel conference and they specifically discussed the nature for this in one of the seminars. It has to do with the fact that we design steel, say for Fy = 50 ksi, but in reality, the steel has a Fy greater than 50.
This 'extra' strength capacity, actually reduces the ductility of system. When an R3 is used, the seismic forces are 'reduced' due to the ductility of the system. Since the ductility of the system is reduced for an R3, you need to be 'add' force back into the design to account for the loss of ductility. This is an oversimplification, but it is basically what I understood from the seminars.
RE: When to and when not to use the overstrength factor. (Structural) 12 Jul 08 01:21.
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